ANStructs

“Most of the resources (books, online forums, videos) I came across on two way slab indicate that moment "will" be higher in the shorter direction than in the longer direction..” In most books, coefficients method for slab design is derived on the assumption that a slab panel is resting on rigid supports i.e. non-deformable supports. Therefore, stiffness of supporting members do not affect the distribution of load from slabs to supporting members. Hence, shorter side of slab which is the more stiffer path carries more load than the longer side. Whereas in a 3D structural model comprising of beam, column and slab, load is distributed based on relative stiffness of elements connecting at a joint. Stiffer element attracts more load and as in your case, the longer slab span proved to be more stiffer load path than the shorter slab span for load to flow. “I want to keep the beam cross section in the long direction as it is, say due to architectural constraints? Is it a hard and fast rule to "make sure" we have more moment in the short direction? Should I always adjust my beam-to-slab stiffness ratio to make sure that happens?” No, you’re not restricted to make sure every shorter direction is the stiffer load path. Assuming the computer model is error free; I would suggest you to properly perform the service and strength design based on the forces you’re getting. “Does it have economic implications?” Not that I know of.

First thing that come to mind is, beams in longer directions are not stiff enough to restrain the deformation of slab. Try increasing the stiffness of long beam and you might start getting the "one way load distribution" as assumed.

Look at page 20 of this document.

Hollow concrete blocks if made following standard procedure are usually very durable and have substantial strength. As far as i know, these blocks have low water absorption than conventional bricks. I think if they are properly placed with high quality mortar, you shouldn't have a problem. You should take my advice with agrain of salt since this type of construction is not very common in pakistan.

Both the gravity and lateral load due to soil on the active side should be considered in the calculation. The soil resistance on the passive side however could be ignored to: 1. Simplify the calculation. 2. There is a possibility that owner of the surrounding property can excavate the soil mass on the other side of the boundary wall in future without your consent. 3. Soil full passive resistance is something that develops as the soil mass is pushed due to significant wall rotation (approximately in order of 0.02 to 0.04 rad). And a designer would not really want this to be true for the serviceable live of the wall. And, apologies for the delayed response.

I don’t believe that modeling the wall with a pin support at ground level resemble the actual condition of the wall. The model more close to reality will be a wall with lateral springs distributed along the embedded depth. This is easier said than done, because, we all know that soil is not a linearly elastic material and calculating its lateral subgrade modulus will not be as easy as dividing the, let’s say, passive resistance with a correlating estimated deflection. And before you ask, No, I will not be of much help in calculating the said property of the soil because most structural engineers (including me) are in a state of blissful ignorance when it comes to the properties of soil. We mostly rely on our geotech counterparts to provide us with all pesky soil related properties, such as, allowable capacity and subgrade reactions. Now as for solution to your problem if I were you, I’ll simply design the wall, for both serviceability and ultimate limit states, ignoring the soil on the passive side and considering the total length upto the base of footing loaded with wind and active soil pressure. This, I believe, will be a safer approach in case of a boundary wall, because, you can never be sure of the possibility if someone is not going to dig up the soil on the other side of the property in future.

Also, for full length of wall and different udl on wall, the formula for displacement will not be applicable.

Yes, your reasoning is right. Calculation steps looks ok. But for calculating displacement, i would take L = 2.5 m since this is the length on which wind udl is acting. If you want to take the full length of wall upto the base of footing, then you need to consider the soil lateral load on the wall as well.

In the computer model, if a designer has assumed that the stairs do not provide any lateral stiffness, then it is the responsibility of the designer to detail the connection such that it reflects the behavior of assumed connection in the model.

Assuming the support of the wall is acting like an ideal fixed support, you can easily determine the deflection at the free end of a cantilever for uniformly distributed load using u=WL^3/8EI. To take into account the cracking of concrete you just need to use the effective moment of inertia in place of gross moment of inertia.

I'll do the following to make sure beam/columns of dual frame can resist atleast 25% of base shear. Find out the %age base shear resisted by column In the original model. Determine the magnitude of base shear that the column should be able to resist, lets say 'x' units based on 25% requirement. Save the original model with another name for design of beam/column only. Increase scale factor of EQ untill the sum of shear force due to EQ of columns at base is equal to x. Design beam columns for relevant load combinations for this magnitude of EQ. Go back to original model and update beam/column design. Check the %age base shear resisted by column in the original model to make sure that columns doesn't start attracting shear force more than 25%. Otherwise revise the design again for higher forces.

Even though you have a higher yield strength of steel, for SMRF, ACI code restrict the value of fy to 60000 psi for axial, flexure, shear and torsion design. Refer to ACI 318-14 Table 20.2.2.4a

Is it only me, or someone else feels that this forum is too quite? Is it usual or is it due to lockdown in some parts of the country? I'm fairly a new member here. Apparently, I joined this forum a long time ago but somehow forgotten about it until recently (I found this site quite accidentally to be honest). I found some very informative posts in this forum. Most post has number of views in 100's, but not many members contribute to the discussion. What's the reason?

In short, a Pakistani PE licence has no benefit or use outside of Pakistan. You'll have to pass the necessary exams and test of that country.

And some people say we will still be dealing with Covid 19 in 2021. 🤔 But, Hopefully not.

May be I'm one of those who see the glass half empty 🙂 . I've become a little (ok, alot) cynical from current situation of Pakistan that I've lost all hope. But that doesn't mean that I'll not encourage or pray for the success of the progressive work done by good samaritans like you. And, I would love to believe this of myself (even though the lack of evidence) that I'll also contribute in such endeavors if I'm in a position to help.

And the discussion has veered off from my original query. Does no one has an answer for the original question posted?

Civil Dept. at NED University have been talking about doing this since I was a student. I don't know where this debate has progressed to since. May be a member from NED civil faculty can shed more light on the status of this project. All good points raised and I applaud you for your endeavors regards to your believes. Nevertheless, IMO the individual efforts will not affect the tip of the iceberg that are our issues. That is, our own construction policies or the lack thereof. We will still be at the same place after a decade or two, until serious efforts are taken at government level. The most immediate step that should be taken is to develop and properly enforce the Pakistan building code and not the ACI code. Even though we all know that the current Pakistani code is the combination of UBC 97 and American codes provisions. Next step will be to immerse the various research work into the code body. That's how we will truly start to have our own code to follow.

For, registered graduate engineers maybe, but not for professional engineer license. I think the problem is not the lack of local or national engineering bodies, but the responsive system of existing governing bodies. The lack of communication channel between these bodies and professional engineers is also one of the reasons we feel isolated from practical world. We are not aware of the constructive work they have done or which might be in progress. In truth, most engineers are satisfied if their 9 to 6 job paid well and do not even care about such matters. There are definitely lack of opportunities in Pakistan for personal growth, to collaborate with peers and share industry best practices. You're talking about your work place dynamics or general observation from this forum? If the former is true, then this is mostly a global phenomenon. There are very few professional engineers who don't hesitate to share knowledge and hard earned experience with juniors.

To those members, who had a PE licence in Pakistan and have now moved to foreign countries. Do you find your PE licence aquired in Pakistan beneficial or useful in other countries?

Practically, the behaviour of a bolted connection resembles to a "pinned" connection more than a welded connection for a connection designed as "pinned". Because welded connection are more rigid than a bolted connection. However in reality, a bolted or even a welded connection designed as a pinned connection, provides some degree of resistance to the moment. Very few connections in reality are truly pinned or truly fixed. And, while a hinge connection could be considered a “worst case scenario” for the beam itself, it does not hold true for the connection, and the supporting member. Because, a "wannabe" hinge connection will in fact transfer "some" moment and it's possible that the connection, or other elements that the connection is attached to may be too weak to resist that moment. Now coming to your question, a gusset plate is a mean to connect a member to a supporting member. A member can be fastened to Gusset plate using either bolts or weld. Direct welding can also be done, and is definitely a more cost effective option but, it could results in field fit-up problems. As to the degree of resistance to moment, If i were you, I would ask following questions to myself. 1. Do I need to be concerned with the possibility of end moments? Depends on how large an end moment I'll get. To check this, I'll perform the analysis for full joint stiffness (i.e., no end releases) and see how "big" the end moment I'm getting. If very small, I might neglect these end moments after cursory making sure it'll not cause any unwanted yeilding in my connection/supporting member. For large magnitude, I'll either provide a fixed connection or will properly model the provided joint stiffnesses. 2. How rigid is the provided connection? Depends on the thickness/size of connection plate and spacing of bolts or weld length/distances. For exact estimation, I believe there's a method specified in euro code with detailed guide lines. (And before you ask, I don't remember the reference off my head, you'll have to surf the net for this one)

I'll add, Both the beams behave similarly in "flexure", for a simply supported beam. For a fixed end beam, special detailing need to be provided between beam and columns to transfer the end moments. For "Shear", however, the behaviour is completely different. Because the "compression struts" carries tension and does not remain in compression, unlike downstand beams. So the shear resistance provided by concrete compression struts becomes close to zero. IMO it's safer to design upstand beams considering Vc=0.

There's an introductory chapter in every code which specifies the use and limitations of that perticular code. I didn't have the above mentioned code. So I downloaded it and as far as i can see it seems like a detailed version (or the application) of ACI 318 and ASCE, combine together, for low rise buildings.

Finally finished reading the debate whose sole participent was SimpleS Structure(just joking. No offense intended).. I would like to add a few comments on what you've said so far. Regarding hinge detail.. All your points seems valid "if" the construction of a hinge detail is carried out like a normal column. However, this is not how a hinge detail is designed and constructed in an rcc member. I'll give you my 2 cents on those of your points which i don't agree with (using your reference points numbering). Although, I don't think this action will be necessary. What's more beneficial than this that this detail will serve it's intended purpose and will not transfer any significant moment? Offcourse, there'll be some limitations such as how much load this can bear or whether this should be provided where large lateral loads are expected. Assuming I'm providing a hinge detail in an RCC column, such as shown in the attached figure, then I'll make sure that the contractor knows about the requirements for vigilant inspection and potential stability issues, and will take preventive measure by providing adequate braces untill the upper floor/beam shuttering is removed and desired load path is achieved. Also, as far as I know, these sections are designed for a minimum accidental eccentric moment due to axial loads. Still, the tensile stresses produced due to these moments are not allowed to exceed the tensile strength of concrete. Well actually, most of the hinges are engineered by locally reducing the size of column. This effectively reduces the moment of inertia of the cross-section which results in very small rotational stiffness and zero (or close to zero) moment from analysis. Yes, the reduced section will cause a very high compressive stress, due to which, a very high compressive strength of concrete and bursting reinforcement may be required. The gaps might or might not be filled later on with an elastic compressible material. As I can see, all these provisions are considered in the picture provided by AliShan. I think the point you're making here is that column will see very little to no moment. And i agree, that's what dual frames are for. But, what if only moment frames are provided to resist lateral loads? By designing for partial fixity condition, you're ensuring that plastic the hinge will form at supports first. Which is not a good design practice and might produce very high P delta moments.

@Ali Shan Short Answer is, Yes. This way you will definitely be on the safer side as far as the design forces for foundation and structure are concerned. As it's already pointed out in previous posts that, the actual behaviour is something between a pinned and fixed support condition. The actual rotational stiffness of the support depends on the interaction of two main factors. 1. Flexibility of the soil. 2. Rigidness of foundation. The higher the flexibility of the soil is, the more it will allow the foundation to rotate (act more like a pinned support) no matter how rigid is the foundation. Similarly, a flexible foundation will not resist any meaningful rotations no matter how rigid is the supporting soil. If you want to learn more about this, try googling "Soil-Structure Interaction".