Syed Umair Haider

Dear All, Kindly share your experience regarding type of support assignment (hinge or fixed) in super structure models for computation of pile loads in case of, a, Columns supported in isolated pile caps. b, Columns supported on piled mats. Emphasizing on following parameters, 1, Difference in Pile loads extracted between both cases. 2, Difference in super structure column design between both cases. 3, Difference in lateral stability of super structure between both cases.

Dear Waseem, Your first two points from NEHRP are logical as they are directed towards the implementation of an upper limit on time period to be used in seismic forces evaluation to eliminate the possibility of excessively flexible structure.These provisions are present in all building codes and seem to be logical and valid. For usage of Ta when Tactual < Ta, The RS curve you have shown indicates Ta & Tc both on constant acceleration zone , so whether you use Ta or Tc you will get the maximum ground acceleration or in ELF terms whether you use any of them equation of max base shear will govern.So both are equal mathematically. However, ASCE endorses what you are saying. But as long as usage of Ta (approximate period) when Tactual < Ta is concerned it still seems illogical to me and will be associated with potential inconsistencies, some of them could be as follows, Drifts evaluated from Ta will be lesser than actual, therefore building separation widths (most importantly) will be under estimated. Code defined equation of max base shear will be of no value for a certain height of structures, while following this. As the actual seismic forces will be greater than that used in design, therefore structure will require high energy dissipation demand than anticipated, as members will be yielded and extended to inelastic range of stresses earlier (due to under estimation of seismic forces) than anticipated stage (associated with R used and detailing performed). Many other issues will also arise due to under estimation of seismic forces that you can also consider. Therefore, although this document is published from a credible source but they didn't justify the adequacy of this provision and didn't address the probable consequences.So it seems a non-engineered dictation to me. Practically this situation is very rare but I will go for UBC97 in such situation rather than following this.

Dear Waseem Your concern is valid but the lower limit given in FEMA example doesn't make sense to me as in a general the usage of time period in seismic analysis is not more than the evaluation of ground acceleration that will be imparted in structure in accordance with specific structural characteristics. Similarly if the specific structural characteristics of a building reveals a time period lesser than generalized approximate time period , then it doesn't make sense to skip the accurate time period and to use approximate one when code itself defines approximate time period as a basis to start analysis for actual T. More interesting is to note that indicated FEMA example is based on 2009 NEHRP seismic provisions , whereas in a separate document issued by NEHRP afterwards (named expanded seismic commentary to ACE-10) it is recommended to use "Tcomp" if "Tcomp < Ta". Moreover, i didn't find any lower bound on time period in UBC97 and even in ASCE 7 which is based on same ELF procedure as given in FEMA example. Therefore, it seems that indicated provisions in FEMA example are overlooked that they have fixed later and usage of Tcomputed if "Tcomputed < Tapproximate" seems valid.

I think both ETABS and UBC 97 provisions are correct. As ETABS completely follows UBC97 therefore only UBC's provisions are elaborated below. UBC seismic design philosophy limits the minimum time period (in the form of maximum base shear ) to Ts (UBC response spectrum) which is equivalent to Cv/2.5Ca. For elaboration , V=CV.I.W/R.T Substitutue T=Ts=Cv/2.5Ca in above eqn therefore V=(CV.I.W/R)X(2.5Ca/Cv) i.e V=2.5Ca.I/w (the maximum base shear that ETABS and UBC97 uses) Therefore time period regardless of its shorter value below Ta cannot be lesser than Ts which is in contrast with UBC97 provisions that are inherent in ETABS. However, analysis indicating time period lesser than Ta indicates a over stiff/over design structure, therefore it is recommended to maintain a natural period closer to 0.1x (no of floors).

W salam, In monolithic construction , T or rec beam is not a matter of choice of designer but its indicated by analysis that if depth of compression block lies below flange (flange = slab thk in beam-slab system) then the compressive force in concrete is balanced by a certain width of flange + total width of web, otherwise only web balance compressive force i.e example of rec section carrying slab load in monolithic construction. Example of rec beam sections also exists in some specific cases of construction for eg RC beams supporting metal deck system,hollow core slab panels and precast construction system. ETABS checks the eqn "a < or > ds" for maximum analyzed Mu-bott and design positive R/F accordingly.

Dear waqas, Calculation of Ie for live load only doesn't make sense as live load will always act on the section after the application of dead loads and will act simultaneously with dead loads. Therefore , Ie must be calculated for possible combinations that could be 1.0D,1.0D+0.5L & 1.0D+1.0L as per ACI-435. Detailed calculations of deflection for 1-way NP flexural members as per ACI 435 is shown in attachment.

In a multi storied building with post tensioned floors, What could be the effective way to model or include secondary (hyper static) forces on vertical supporting elements i.e columns & walls due to post tensioning, that seems to be typically includes inplane diaphragm force (P) & moment due to eccentricity of tendons to cross sectional centroid of slab (P.e). Kindly share your views and practical approach towards it.

Auto meshing doesn't ensure adequate connectivity between member to member and is therefore recommended for horizontal area elements enclosed by line elements only (floors), where there is no structural connection between floor and any element in between the panel. In case of vertical elements, connectivity between vertical and horizontal elements is of due importance and is better to be achieved through manual meshing. In case of auto meshing as you indicated, change in size of auto mesh could solve the problem as its possible that connections inadequate (nodes not coinciding) on 1m element size can be adequate for 1.2m size (nodes start coinciding) and so on. For p-delta,a possibility exists that due to any meshing error some connection is modelled with inadequate lateral stiffness i.e when program try to impose lateral deflection due to seismic loads,modes start yielding frequency below shift. If you are interested in studying the problem,then easy approach is to check each mode shape and investigate the member that is going in unrealistically large displacement. Solving this member's connectivity inadequacy will solve your problem.

This message is indicative of some modelling error in your model as Shift is the centre of cyclic frequency range that is used to limit the modal frequency range i.e to enforce program to neglect modes with frequencies below shift,By default shift is a very small value something like .0001,any mode with eigen value lesser than that of shift means a mode with a very low frequency or a very high modal period that indicates some modelling most probably member connectivity error in your model.

Yes you seem to be correct.I am not familiar with ldh restriction in zone 4, I will check it.

Regarding point # 3, suppose at any support strength requirement shows 3 #6 bars,here to counteract ldh you provide greater area i.e greater bars for ex if you provide 6 # 6 bars then ldh should be multiplied with As-3#6/As-6#6 i.e 0.5. Concept behind this provision is that if at any support say half ldh is available for stressed bars then these bars will develop half of their strength through bonding in concrete and will be able to take stresses till 0.5xFy,if you add up same no of bars with same ldh then they will also bear stresses upto 0.5xFy and as a whole you will get the required resistance Fy.

Waqas, First note that development length in indicated case is only the straight length as shown in figure and doesn't include bent portion and secondly this requirement can be fulfilled as follows, 1,you can use greater no of lesser dia bars complying ldh requirement and if beam width is insufficient to accommodate greater no of bars then you can distribute them in effective flange width (ACI 318-11 section 10.6.6). 2. you can multiply ldh by 0.8 if you tie hooked end reinforcing bar with ties parallel or perpendicular to the bar, spaced not greater than 2db (db = hooked bar diametre).(ACI 318-11 figure R12.5.3). 3, You can provide excessive reinforcement and multiply ldh by As req/As prov. 4, You can multiply ldh by 0.7 by maintaining minimum side cover to the longitudinal bar to 2.5".

Uzair, See chapter "foundation structures" topic foundation structures for frames.

Asad, It is not necessarily required to extend the seismic analysis's storey range till basement levels as for buildings with several below grade levels supported by basement walls, two stage static analysis procedure is used (ASCE 7-10 Section 12 & UBC97 section 1630.4.2) that consists in distribution of building in flexible upper portion (above basement levels) and rigid lower portion (basement levels), provided the lower portion have a stiffness minimum 10 times greater than upper and time period of whole structure should not exceed 1.1 times of flexible upper portion's period while it is considered as a separate structure. You can simply check these limitations as, 1, by computing stiffness ratio (EI/L ratios of basement walls + LFRS in rigid lower portion) to the (EI/L ratios of LFRS in flexible upper portion) 2, computing time period of whole structure (Eigen vector) and computing time period of upper portion alone modeled without basement levels. Having satisfied these, seismic analysis is required to be performed till base of upper portion only & rigid lower portion is required to design only for seismic forces transmitted at the base of flexible upper portion modified by the factor Rupper/Rlower. In ETABS you have to define "ground level" as bottom storey in analysis storey range and seismic shear imparted on ground level will be automatically transmitted to the levels below through diaphragm action.It will be just required to compute "R" value for lower portion considering it separate and to modify seismic load case's scale factor by Ru/Rl for the design of below grade structure. In this way the maximum seismic shear will be acting at the ground level not at B4, that will reduce the magnitude of force and could be beneficial in mentioned below grade serviceability issues particularly drift will be considerably reduced (also compute drift using user defined time period obtained from eigen vector analysis see UBC Section 1630.10.3). As long as below grade torsion is concerned, it is just required to satisfy that Ax (UBC97 Eqn 30-16) should not exceed 3 and required to be noted that amplification of diaphragm eccentricity is of no meaning there since seismic forces are imposed from upper portion and are not calculated & applied separately. Secondly, load combinations should be inclusive of minimum seismic vertical effects and dynamic load combinations.

It is applicable to SMRF i.e beams & columns and structural walls with their coupling beams.In general all members effective in resisting lateral force.

As Umar said the real response would stand somewhere in between a pin and fixed base, similarly extent of column fixity at base depends upon the rotational flexibility of foundation that depends on foundation stiffness and soil stiffness both. In general it can be said that for rigid raft foundations,foundation supported on stiff piles or basement walls this rotational stiffness is high and one can confidently assume fixity of column at base. But in conditions of individual footing pads on deformable soil,foundation could have considerable rotational flexibility and consequently assumption of fixed column base could show a considerable variance with respect to distribution of column moments on bottom storey i.e in such case column moments could be concentrated at top end of bottom most column rather than at base. Therefore, It is recommended to model the rotational stiffness of foundation rather than assuming fixity at base in cases of individual column pads on deformable soil to represent quite realistic partial fixity available at column base.This methodology can be seen in "Seismic design of reinforced concrete and masonary building by M.J.N Priestley" (page 466).

W salam, These concepts need some fundamental theory of development of seismic analysis procedures (static procedures) in codes as briefed below, When engineers decided to go for an earth quake resistant design,then they initially proposed to assign a horizontal load of "0.1 x Weight of structure" to cater for seismic forces.With the passage of time several geo-technical and site specific response characteristics were included in analysis for evaluation of seismic forces and structural members were designed to resist these forces in their elastic range i.e to not yield under these forces. Structures designed accordingly surprised engineers, as they were observed to show little tolerable non structural damages in seismic events considerably greater then those considered in evaluation of seismic forces.This leads to the development of concept of energy dissipation and over strength factor i.e under cyclic seismic loading structures have the ability of resistance beyond the elastic range of stresses in members (after yield), in proportion to their ductility.Since then started consideration of this structural over strength characteristics that consists in reduction of design seismic forces in accordance with their energy dissipation characteristics or mathematically reduction of base shear by division with over strength factor. For ex in accordance with UBC97, if on a structure the actual seismic force i.e Cv.I/W=550 & over strength or ductility factor is R=5.5 then Seismic shear will be 550/5.5 = 100, then structural members will be designed to remain elastic or not yield under the lateral force of 100, whereas they will dissipate the remaining 450 in inelastic range or in terms of energy it can be said that this structure is able to dissipate 450/550x100 = 81% seismic forces through its ductility and is required to design elastic only for 19% of actual seismic forces. In the lights of above these clauses could be defined as follows, ") 21.1.1 says, ........................................For which, design forces , related to earth quack forces, have been determined on the bases of ENERGY DISSIPATION IN NONLINEAR RANGE OF RESPONSE". For every structure,seismic forces are evaluated in accordance with corresponding over strength factor that indicates the extent of probable energy dissipation. 2) Commentary of R 21.1.1 says, The integrity of the structure in the inelastic range of response should be maintained because the design earth quack forces, defined in documents such as ASCE/SEI 7, the IBC, the UBC and NEHRP provisions are considered less than those corresponding to linear response at the anticipated earthquack intensity. As defined above, structures are designed for seismic forces that are reduced by over strength factor however actual fores are times greater than that, therefore code requires that when structure is subjected to actual seismic forces(plastic state),then although structural damages in members are tolerable but integrity of structural members should necessarily be maintained so that structure will not collapse.This condition is another form of philosophy of safety in code under seismic events that says "Under major earth quake,structure should be designed to have structural & non structural damages but should not collapse".

Dear Zain, I don't know the origin of document,you have uploaded for calculating torsional constant,but the methodology given therein is incorrect.As "Tcr" and "Tu" given therein are indeed threshold torsional strength and ultimate torsional stresses respectively, and are both design properties not analysis properties. (See ACI 318-11 section 11.5.1). Whereas the torsional constant, ETABS asks in "analysis property modification factors" is simply the torsional moment of inertia (J) used to determine torsional stiffness of a member (JG/L) i.e something else. As long as its value is concerned,then in building structures it is a general practice to use a negligible value like .001 to nullify beam's torsional stiffness.In this way, the torsional stresses (if arising due to compatibility of deformation i.e compatibility torsion ) are transferred via alternate load path (i.e redistribution of torsional moments occurred), considering that beam is unable to provide torsional restraint and in other condition if torsional stresses in beam is required to satisfy equilibrium of structure (where redistribution is not possible) then torsional stresses in beams remains independent of whatever value of "J" you have selected as equilibrium equations are necessarily satisfied independent of stiffness as "Compatibility is optional and equilibrium is essential". This approach of minimization of "J" economize beam sizes that arise from stringent combined shear and torsion requirement of building codes,but consequently beam sections designed in this way will start developing internal horizontal cracks (hairline cracks not affecting functionality of structure) due to torsional stresses and their torsional strength will continuously degrade till the design condition is achieved i.e negligible torsional strength of beam.But as the structure is designed to be stable without torsional stiffness of beam so it remain stable after this condition is achieved.However, the beam member itself cracks that doesn't affect the functionality of structure in any way. A very descriptive and clarifying description is available in "Reinforced concrete design by Arthur Nilson". As long as authentication of this approach is concerned then it is allowed by building codes as, 1, ACI-318-11 section 11.5.2.1 & 11.5.2.2. 2, UBC97 section 1911.6.2.1 & 1911.6.2.2 3, BS 8110-1 1997 section 3.4.5.13 Keeping in view above mentioned, it is a general practice to nullify torsional constant of beams in building structures and it is not required to use any iterative process to derive torsional constant of each beam section that is indeed not practical as there will be thousands of beam span in large structures.

W salam, UBC97 Section 1612.1 states "Buildings and other structures and all portions thereof shall be designed to resist the load combinations specified in Section 1612.2 or 1612.3 and, where required by Chapter 16, Division IV, or Chapters 18 through 23, the special seismic load combinations of Section 1612.4". Hence it seems that these combinations don't intend for general design usage & should be only used where specifically required by code. Further the requirement of special seismic load combinations is specified in code in only two provisions i.e "In design of elements supporting discontinuous members and collector elements, in sections 1633.8.2.1 and 1633.2.6 respectively" as follows, "1630.8.2.1 General. Where any portion of the lateral-loadresisting system is discontinuous, such as for vertical irregularity Type 4 in Table 16-L or plan irregularity Type 4 in Table 16-M, concrete, masonry, steel and wood elements supporting such discontinuous systems shall have the design strength to resist the combination loads resulting from the special seismic load combinations of Section 1612.4." And, "1633.2.6 Collector elements. Collector elements shall be provided that are capable of transferring the seismic forces originating in other portions of the structure to the element providing the resistance to those forces. Collector elements, splices and their connections to resisting elements shall resist the forces determined in accordance with Formula (33-1). In addition, collector elements, splices, and their connections to resisting elements shall have the design strength to resist the combined loads resulting from the special seismic load of Section 1612.4." Except these, no where in the code, these special load combinations are advised to be used.Therefore, allowable stress design combinations for your intended stresses evaluation should be independent of these special combos.

"Assuming the diaphragm as rigid, there will be negligible in plane displacement of diaphragm with respect to its supports,which are vertical elements of lateral load resisting system, at particular story-level. Lateral displacement of diaphragms is with respect to story-levels. Hence, the support provided by slab to basement wall, at a particular story-level, can be considered as hinge as it will resist translational movement due to in-plane stiffness of rigid diaphragm." All written above regarding rigid diaphragm is correct,but in fact I am indicating the net displacement of diaphragm along with connecting LFRM rather than its in plane deformation relative to LFRM which is assumed as zero in case of rigid diaphragm. For eg if we suppose six basements of 12' each and assume that inter storey drift in each basement is 1" then net displacement of diaphragms w.r.t origin at 5th,4th,3rd,2nd & 1st level will be 1",2",3",4" & 5" respectively and as these basement slabs are acting as a horizontal support to basement walls therefore the support deflection @ wall's 1st,2nd,3rd,4th & 5th support will also be 1",2",3",4" & 5" i.e these slabs will not be acting as a rigid horizontal support to basement wall at storey levels but instead they will be undergoing lateral deflection along with wall.In this case the deflected shape and correspondingly flexural behaviour of wall will be quite different from the case of propped canti lever model that enforces assumption of basement slabs as rigid support with zero horizontal displacement.While propp canti lever assumption works adequately for the case of entirely confined basements from all sides but will not simulate the condition defined here due to absence of lateral confinement from other side. I have modeled both cases separately and have attached herewith deflected shape diagram extracted from both cases that indicates the difference in their behaviour and correspondingly in flexural forces. Why would someone be worried about out of plane displacement of basement wall, when it will be proportioned for out of plane forces( earth pressure). As it has been correctly been pointed out that drift limits in building codes are intended to control NON/STRUCTURAL damage, and to limit secondary forces due to P-delta effect. The point is: one should be worried about drift of frame (in/plane relative displacement), rather than the out of plane displacement of basement wall. It is correct that basement wall's out of plane bending is not a sensitive issue when it is proportioned for out of plane forces as you have indicated.Similarly all indicated above is about its correct proportioning w.r.t condition defined by Rana (significant lateral drift in below grade levels). I doubt if you can do that. The reason is that basement wall backfill would least likely to be compacted. Imagine doing that would increase load >> significantly which is not desired. Normally the lateral spring coefficients provided by geotech report are for undisturbed soil or soil well compacted. Your doubt is meaningful for shallow depth basements but I think in case of several below grade levels, soil bed could be reasonably compacted by overburden pressure from above lying soil itself as an example overburden pressure @ depth of 40' could be 4800 psf for a soil density of 120 pcf. This high overburden generally enables soil to create a massive passive resistance. For an eg passive resistance of soil for a basement wall of 200' length at same depth (in cohesion less soil) will be as follows, (Brinch Hansen's eqn), P=3.H.B.Kp.Density=3.(40)(200)(3)(120)=8640 Kipps that seems to be a considerable value. Dumped backfills would provide much softer springs that I would suggest be considered as void regions (Same thing as designing a pile for lateral). Above mentioned cases demand sound engineering judgement for consideration\unconsideration of lateral soil support but if loose characteristics of adjacent soil is certain,then all below grade levels (regardless of single\all side's confinement) should be included in seismic analysis and base location for seismic shear must be at foundation level that automatically includes all below grade levels in drift analysis without any side support. Thanks & Regards.

Below grade seismic drift consideration. Considering building structure having several below grade levels and basement walls on a single (or two) sides, I like to share my views as follows, First of all drift limit states are intended to limit the damages in non structural elements in case of lateral movement and from structural point of view to limit the second order effects to a tolerable extent. Therefore, drift indexes defined in code are subjected to diaphragms rather than individual structural elements (like basement walls).It is advised to check serviceability considerations in accordance with diaphragm's drift limitation of code. As long as underground structural drift analysis with single side confined is concerned,two separate cases that seems to be required for consideration are as follows, 1, When structure moves towards the fill (passive condition). In this cases due to infinite stiffness of soil (assumption to be verified as defined later) adjacent to basement wall,displacement of soil will be negligible and correspondingly differential displacement of basement wall from bottom to top will also be negligible,that creates almost zero drift in below grade floors (diaphragms) if they are rigidly connected to basement wall (i.e the case of monolithic slab-wall joint with adequate reinforcement anchorage for tension directed in plane of slab). Modelling Strategy: This case can be simulated in model by assigning relevant basement wall throughout its height with compression only spring where spring represents lateral soil support.(Usually lateral soil support is modeled using P-Y curve).Then an application of seismic force in the direction towards soil will yield a realistic value of lateral deflection of soil (could violate infinite stiffness assumption) incorporating the effect of lateral soil support. In this case considerations must be given to the design of basement wall as it will be experiencing different earth pressure profile due to seismic excitation.Therefore it must be designed for this (seismic) pressure condition as well. 2, When structure moves away from the fill (active condition). In this case as there is no wall to confine structure on the other end,therefore the net drift will be Seismic drift + lateral earth pressure's displacement.Both of these forces will be unidirectional in this case and will be effective in diaphragm drift. Combined drift can be evaluated using loads combination from UBC97 section 1612.3.7 with exception 1612.3.3 (note that there "H" indicates lateral earth pressure). As seismic drifts are required to be amplified to represent inelastic effects then there amplified values must be used in above mentioned serviceability load combinations (could be done through scale factors) whereas drift due to lateral earth pressure doesn't need to be amplified. "2) I still did not get why there would not be any problem in basement for drift. I have multi-basements. Do you mean there would not be excessive drift between basement stories? I am getting it and it is more than allowable seismic drifts....guidance needed pls!" If this is the case i.e you have considerable inter-storey drift then an important consideration here must be the thickness of basement wall.In this case the support condition hinge (i.e zero horizontal displacement) cannot be assumed for basement walls at slab levels as diaphragms are undergoing considerable lateral displacement.Therefore, in order to evaluate flexure in basement walls it is advised to model a line element along the height of wall (meshed at wall area edges) to compute flexure in basement walls rather than using theoretical model of prop canti-lever assumption that seems to be invalid due to diaphragm's horizontal movement . It is recommended to design basement walls first (before analysis for drift) based on above methodology and model its accurate thickness, as the in this case basement wall is serving as a tie member between underground diaphragms and its thickness will effect differential diaphragm movement between different levels. "Sir but the point is that slab will distribute the lateral load to all supports. Like seismic or wind. Lateral load is a lateral load and that would keep the earth pressure drift to minimal." I think that lateral earth pressure for several below grade levels could be significant enough to be considered. For eg if we consider 6 below grade levels 6@10' with length of basement wall=200' and Ca=0.33 + soil density=120 pcf then net horizontal diaphragm force due to lateral earth pressure in bottom most (5th diaphragm) above foundation will be as, p(intensity@mid height b/w 5th& 6th level)=Ca.We.h=(0.33)(120)(55)=2178 lbs/ft2 p(intensity@mid height b/w 4th& 5th level)=Ca.We.h=(0.33)(120)(45)=1782 lbs/ft2 Force (tributary to 5th diaphragm per unit length of basement wall)=(1782x10)+0.5x(2178-1782)(10)=19,800 lbs/ft Net force tributary to diaphragm=200x19,800=3960 Kipps It seems to be a significant figure but needs to be checked as in contrast with magnitude of seismic force at same level. Thanks

.003

1, 40 psf can be said a high value with regard to my experienced design snow load for similar nature projects in various localities of Pakistan and it is confirmed by Hasan as he has revised it to 22 psf after incorporating several factors. Moreoever,in Hasan's case usage of MBMA 1996 (Metal Building Manufacturer's Association) provisions for snow load calculation,reduction & application in contrast to truss system, should be more appropriate rather than ASCE as MBMA adresses specifically such types of structures. Moreover,as you recommended to assess snow loads in Pakistan using ASCE/SEI,how can you calculate Ground snow load "Pg" for your area as specified in section 7.2 which is based on American localities only?. Conclusively,if you intend to use snow loads in Pakistan you have to go for several assumptions due to unavailability of precise information and in hasan's case (steel structure) appropriate assumptions could be made using MBMA 1996 that will differentiate ASCE mainly in terms of application of load. 2, Yes it is evident that couple could only exist in case of bott & top chord connected to side face of beam only.In case of single chord connection torsion could only exist if the chord is provided with a moment connection in supporting beam that is generally avoided for truss structures.

40 psf is a high value for snow load that could result in an uneconomical truss design.You could check for snow load reduction provision (UBC97 Eqn 14-1) if not applied. Moreover structural beams that will be used to support truss will necessarily required to be cater for torsional design as they will be subjected to high torsional moments in two ways, 1, Axial forces in bott & top chord of truss will create a couple acting about the longitudinal axis of beam i.e creating torsion equivalent to axial force times depth of truss. 2, While providing slotted holes (roller supp) to eliminate thermal stresses you have to embed a bracket plate into beam that will be supporting truss and correspondingly will be subjected to vertically directed pressure (from truss end shear) acting at an eccentricity equals to 0.5xbeam width + distance of beam edge to centroid of loaded plate area i.e creating additional torsion moment in beams equals to truss's end shear times eccen.

It seems that ACI is clear in its provision and usage of 0.6% rather than 0.2% reinforcing ratio for temperature & shrinkage crack/stress control could not be justified for general structures and is only applicable in cases where crack width is a design consideration. As ACI -350 table 7.12.2.1 recommends only for "structures subjected to environment or required to be liquid tight" a reinforcing ratio range from .003 to .006 for Grade 40 & .003 to .005 for Grade 60 steel as a function of length between movement joints from 20' to 40'.(see ACI 350 7.12.2.1). Thus the extreme ratios ".005" & ".006" represents the maximum temp & shrinkage reinf ratio for "structures subjected to environment or required to be liquid tight" where movement joint is not provided.This maximum ratio is mentioned in ACI 224 and is supposed to be followed when crack width is a design consideration for eg in structures as defined above.(note here structures subjected to environment could simply refer to outdoor structures in accordance with CIRIA technical note 107 section 2) On the other hand as long as general structures is concerned (say building structures) crack width is not a design consideration in accordance with same code (ACI 318) and here provision of minimum reinf (0.18-0.2%) intends to limit shrinkage & temperature stresses rather than crack widths within tolerable limits (see ACI 318-11 7.12.1) which is claimed to be observed for many years satisfactorily controlled with these ratios. As long as design consideration of shrinkage & temperature effects is concerned in general structures, we include the effects of shrinkage & temp cracks indirectly in serviceability limit states as 1, In the form of additional vertical deflection due to creep & shrinkage of structural members (see ACI 435R-95 table 2.1). 2, Effective stiffness (see ACI 318-11 10.10.4) to be used in the analysis of lateral deflection accommodates the stiffness reduction due to shrinkage & temp cracking. 3, ACI 318 also recommends to analyze for the severity of stresses arising from temp & shrinkage effects (see ACI 318-11 8.2.4 & 9.2.3) Hence, usage of 0.18-0.2% reinf and 0.3-0.6% reinf is justified for general and special structures respectively.