Badar (BAZ)

Looking at the results of mid-span and at interior support, a difference of 2% in results is not "significant" at all. You have not done anything wrong.

That is a well put response. On which forum did you get this response from?

You should make the judgment about the degree of rotational restraint available at the ends which can be based on calculations, intuition or experience. It depends on the end conditions: single span with both ends of beam supported on masonry >>> no rotational restraint beam's end is supported on RCC columns>>> rotational restraint is available and depends on the relative stiffness of column to beam; one can use software, moment distribution method (MDM), or experience to get the distribution of moment between support and midspan. beam's end is continuous over support, meaning there is another beam on adjacent span>>> rotational restraint is available and depends on the span and depth of adjacent beam; again use software, MDM or moment coefficients from ACI or other similar source for moment coefficients.

Yes, the thickness of wall is appropriate as per your results; you need to provide the prescribed confinement at the ends of wall where compressive stresses are exceeding the limit set by the code.

ASCE_7 refers to ACI 318 for design of reinforced concrete. Strong column-weak beam provision is present in ACI 318, and yes it does need to be more than beam's capacity for structures located in regions which require high ductility against the seismic actions.

You got the wrong interpretation of the concept. These two values serve as reference points to define the shape of response spectrum. Two approaches have been mentioned in the code for calculating the time period. One is empirical, and the other is analytical.

There is no code requirement for it. You will need to make sure that enough length is available for the development of reinforcement at the critical x-section. Also insure the rotational stability keeping statics in mind, if it is critical.

I am not able to understand how can there be a demand of 1380kN axial force on this beam. Any way, you will design this member for combined axial+flexural actions, meaning it should be treated as column. Draw interaction diagram, and see if Pu, Mu falls within the envelope. it depends on the demand-to capacity ratio.

You got them mixed up. They serve different purpose. Stiffness modifier--------> effect of cracked properties at service/design loads deflection amplification factor --------> kind of extrapolates elastic deformation to inelastic deformation for a structural system of known "R" value.

If the structure is design as per IS code, you should follow the guidelines of the code, unless you can prove through a more rigorous analysis that the member will not achieve limit state intended by the code for prescribing this limitation on the spacing. I believe it is related to preventing the buckling of bar and to ensure the confinement of column core.

You can design your members for the forces that you got by assigning the code prescribed stiffness modifiers. It wouldn't be an unreasonable approach since members will experience cracks at the ultimate limit state. But if you decide to tread an conservative approach, and design your beams based on gross cross-section, and columns based on cracked properties, it is still reasonable. But, the later approach will require you to run two more models. one more beams and the other for columns. And if you have shear walls, then you might need another model.

Fundamentally speaking, there is nothing wrong with autogenerated load combinations, provided you know the what is the difference. It is not at all difficult to figure out what will be the difference (load combos have just one variable, some factor x specified loads), just read the relevant code. Your foundation will have a certain value of flexural and shear capacity measured in terms of k-ft and kipps or whatever the system of units you are using at different locations. Why is it difficult to decide what to do with the results of Etabs? You need to compare the results of all ultimate limit state load combinations with the capacity at these locations. You can use different codes within the same family of codes. But, you should keep in mind that the definition of DBE level earthquake is different in UBC 97 as compared to the ones that arrived after it. Having said that, the design methodology (strength reduction factored and loads factors) is the same.

Sorry, there was a mistake in conveying my idea. I have corrected my post.

Yes, I agree. It should be used for wind only. Procedure of ACI's 9.5.2.3 (318-11) should be used for floor slabs as you must include the effect of creep and shrinkage as well. Calculation of the deflections with un-cracked properties only will result in unconservative results for floors. There is no one fits the all answer for this. Stiffness modifiers does not matter for structures under gravity loads. Axial load will be the same for various combinations of stiffness's. Moments can vary, but not much. Moments results from relative stiffness of connecting member. If you will increase the stiffness of beams by some ratio to account for service conditions, then you should increase the stiffness of columns by the same ratio, and end up getting the same ratio for relative stiffness of beams and columns. But for seismic forces, the time period effects the design forces, and depending upon the location of structure on response spectrum, the forces can reduce or increase especially under the core area. You will need to take appropriate action. If a building is more than 15 stores in height, you might end up using the same reduction factors for both types of analysis.

There are various resources. Search for masonry or reinforced masonry and you will find books and design guidelines. I have used US and Canadian code for masonry. I have also used a book regarding seismic design of masonry from Thomas Paulay and M.J.N Priestley as a reference as well. ACI 530/530.1-13, “Building Code Requirements and Specification for Masonry Structures and Companion Commentaries"

We cannot design masonry in ETABs because the software does not have Masonry design code in its library. It will not perform calculations required for determining shear strength of the x-section modelled in ETABS, and it will not check if the compressive/tensile stresses are within limits. But, you can analyze it in the software. Get the forces, and perform the required checks manually.

Yes, you can fill the entire section of minaret at particular height with reinforced concrete and call it a "shear wall". However, It is inappropriate to call it a shear wall, as the term is used in the building frame system where floor, acting as diaphragm, distributes lateral loads to various sparsely spaced vertical members. Calling it a tube, or pipe will be a better terminology. I have seen designs where they did not fill the entire section with reinforced concrete. They only filled the regions at bends with RC and connected those vertical members at regular intervals of 10-15 ft along the height with horizontal RC member which serves to tie them and makes them act as a group. The voids were filled with masonry. If you are dealing with seismic loads, it better to make it all RCC; falling bricks can become a life hazard, unless you make them attach the bricks with RC through some connector.

These equations are not not meant for one-way floor system. You should know it.

How did you come to the conclusion that you need 21" thick slab for one way slab of 26 ft span? Anyone who is not familiar with the project, is not in good position to recommend you to go for two way floor system. Architectural constraints can control. Check seismic demand, and the related ductility demand; the seismic considerations may govern as well.

Without shear studs, horizontal shear can not transfer between RC slab and metal and there is no composite action between metal deck and RC slab . Some manufacturers of metal deck have designed their product such that their decks develops some degree of composite action without shear studs. Depending on the design loads, that may be enough. Check your design loads and related effects to decide if you need 100% composite action.

If you are taking about adding the deflection due to lateral pressure of soil above the footing pad, it will have negligible contribution. I am not aware of it.

The reason is all apparent: Demand was more than capacity. The beam at front failed in flexure as well as shear. Floors collapsed as the section of the column just above the beam-column joint could no longer take excessive demand of moment. It probably had inadequate lap splices as well. May be beam-column joint failed in shear as well.

Sorry for the late reply. Procedure is correct.

Yes, " nonlinear cracked" is ok. A part from that, you need to specify the source of reinforcement for cracked analysis in SAFE through "cracking analysis options". How are you performing your manual calculations?

The maximum seismic force that a structure can experience depends on its mass, stiffness, and damping properties.